Statnamic testing of piles in clay 2

Statnamic testing of piles in clay

It was found that the quake values of the pile shaft resistances of the rapid load tests were much higher than those of the static load tests An existing theoretical model will be modified and applied for the static load pile tests to build the relationship between load and settlement and to quantify these quake values, and then it will be developed for the rapid load pile tests

The gradual decrease of the pile shaft resistance after its peak value to a residual pile shaft resistance, which is known as the softening effect, plus the changes of pore water pressures and the inertial behaviour of the soil around the pile will be reported and discussed

6.2 Typical results of the pile load tests

Several measurements for each test were obtained from the instrumented model pile and clay bed instruments The following definitions of the testing components will be used:

♦ Total pile load - load applied to the pile top and measured by a load cell mounted at the pile top

♦ Measured pile shaft load - load measured by the shaft load cell ♦ Pile tip load - load measured by the pile tip load cell

♦ Total pile shaft load It was not possible to measure the total pile shaft load directly Therefore, it was deduced by subtracting the pile tip load from the total pile load ♦ Pile settlement - vertical displacement from its original pre-load test position measured by an LVDT mounted at the pile top

♦ Pile velocity - deduced from the pile settlement with time or from an accelerometer which was incorporated in the pile

♦ Pore water pressures at the pile tip and on the pile shaft - measured by the pile tip and pile shaft pore water pressure transducers

♦ Pore water pressures in the clay beds - measured by pore water pressure transducers which were incorporated in clay beds at different locations

♦ Soil accelerations in the clay beds at different locations - measured by accelerometers located at different locations in the clay beds

♦ Top and side chamber pressures - measured by Druck PDCR 810 water pressure transducers

Typical results of pile load tests are shown in Figure 6.1 to 6.4 In the following sections these results will be discussed and analysed in more detail

6.3 Pile shaft load results and models for the pile shaft load

The total pile shaft load which was deduced from the total pile load and the pile tip load was less reliable than the measured pile shaft load as the pile tip load cell worked unreliably (see Section 6.4) Therefore, the measured pile shaft load acting on the shaft load cell rather than the total pile shaft load will be presented and used in this section

With a CRP test, the rate of shearing was not constant but it increased gradually from zero to the target rate (Figure 6.2) Normally, the desired rate for a CRP test was only achieved when the pile shaft load had reached the ultimate load value Therefore, the rates of shearing, which were obtained from measured pile settlements, will be used in deriving the pile shaft static load from the pile shaft rapid load In reality a CRP test at a high rate is similar to the first part of a statnamic test

Smith (1960) worked on the dynamic resistance of a pile and proposed a linear dependence of the damping resistance upon the shearing rate However, the nature of the non-linear relationship between the damping resistance and the shearing rate cannot be ignored For this reason, Gibson and Coyle (1968) carried out triaxial tests at different rates of shearing for both sand and clay and proposed a non-linear damping resistance Following this, several researchers have worked on this problem and proposed several soil models for the relationship between the damping resistance and the rate of shearing

In order to examine the capability of the non-linear damping models several typical models will be used for analysis of the test results Following this a new model will be proposed to derive the load-settlement curve of a static pile load test from a statnamic pile load test

Rs is the static resistance JT is the damping factor v is the velocity of shearing

N is the parameter drawn from the test results, which was 0.18 for clays and 0.2 for sands

♦ Randolph and Deeks’ modified model (Hyde et al 2000):

(6.2) where τd is the dynamic shear resistance

τs is the static shear resistance

vo is the reference velocity (taken for convenience as 1 m/s)

Δv is the relative velocity between the pile and the adjacent soil α, β are the damping coefficients

♦ Balderas-Meca’s model (2004):

(6.3) where τd is the dynamic shear resistance

τs is the static shear resistance

vo is the reference velocity (taken for convenience as 1 m/s) v is the relative velocity between the pile and the adjacent soil

α is the damping coefficient It is a function of the pile displacement and varies linearly from zero when the pile displacement is zero to 0.9 when the pile displacement is 1% of the pile diameter After the pile settlement achieves 1% of the pile diameter the α parameter remains constant

Due to the dependence of the damping parameters upon soil properties two sets of damping parameters for each model will be used for the examination of the pile test data The first set of damping parameters are those proposed by the model’s authors, and the second set is deduced from the best match between the test data and the models Thus, the two sets of the damping parameters for three models will be determined as follows:

♦ Gibson and Coyle’s model: These authors only recommended N = 0.18 for clay meaning only one set of damping parameters is applied in the model, with the JT value achieved by a best match between the test data and the model

♦ Randolph & Deeks’ modified model: α = 1 and β = 0.2 are used for the first set of damping parameters and β = 0.2 and α achieved by the best match process are the second set of damping parameters for the model

♦ Balderas-Meca’s model: β = 0.2 and α increasing linearly from zero when the pile displacement is zero to αmax = 0.9 when the pile displacement is 1% of the pile diameter are used for the first set of the damping parameters β = 0.2 and αmaxobtained from the best match process are used for the second set of damping parameters Thus, β = 0.2 and α increasing linearly from zero when the pile displacement is zero to αmax when the pile displacement is 1% of the pile diameter are

the second set of the damping parameters After the pile displacement reaches 1% of the pile diameter α is constant and equals αmax

It was desirable to produce highly repeatable clay beds However, test results showed that the five clay beds exhibited slightly different damping characteristics To demonstrate this two CRP tests at rates of 0.01mm/s and 100mm/s in each bed are compared in Figure 6.5 to 6.9 The ratios of the maximum measured shaft loads of the CRP test at a rate of 100mm/s to that of the CRP test at a rate of 0.01mm/s are 1.97, 1.54, 1.82, 1.74, and 1.95 for clay Bed 1 to Bed 5 respectively Taking the average value of five ratios, 1.8, as a benchmark, the deviations of the ratios from the benchmark are: 9.30% for Bed1; 14.56% for Bed 2; 0.77% for Bed 3; 3.32% for Bed 4; and 7.81% for Bed 5 Due to this variation the matching process between the models and the test results to deduce the damping parameters will be carried out independently for each bed

The pile bearing capacities in constant rate of penetration tests at the rate of 0.01mm/s are taken as the pile static bearing capacity benchmark and are used for comparison with the derived pile static bearing capacities However, the pile static bearing capacity benchmark for each clay bed was not constant since the consolidation was maintained during the testing programmes In addition, significant local consolidation developed on the soil around the pile under a series of pile load tests For this reason, the CRP test at a rate of 0.01mm/s was repeated several times in each bed and a rapid load pile test used the nearest CRP test at rate of 0.01mm/s as its static bearing capacity benchmark

A rapid load pile test and a CRP test at the rate of 0.01mm/s in each clay bed will be used to check the suitability of the existing models The static load-settlement curves, which are derived from the dynamic load-settlement curves by using the chosen models, will be compared with the measured static load-settlement curves which are the results of CRP tests at the rate of 0.01mm/s

The comparison of some load-settlement curves are shown in Figures 6.10 to 6.24 Combining the results of all pile load tests in which some are shown in Figures 6.10 to 6.24 the following conclusions can be deduced:

♦ Rate effects are present and the damping resistance is non-linear with the rate of shearing as suggested by Hyde et al (2000)

♦ The quake of the pile shaft load, i.e the penetration at which the pile shaft load reaches the ultimate resistance, in a dynamic pile load test is larger than that of a static pile load test The quakes in static pile load tests for pile shaft load vary between 0.5% to 0.9% of the pile diameter (70mm), whereas the quakes of dynamic tests vary over a wide range and the quicker the loading rate the larger the quake value The largest pile shaft quake for rapid load pile tests is 5.4% of the pile diameter (B4/12/CRP-400) ♦ The pile shaft load-settlement curve of a rapid load pile test can be subdivided into three sections as shown in Figure 6.21: i) in the first section the relationship between the pile shaft load and the pile settlement was approximately linear; ii) in the second section the relationship between the pile shaft load and the corresponding pile settlement was non-linear; iii) finally, in the third the pile shaft load reached the ultimate value and remained approximately constant with further pile settlement On the other hand the pile shaft load-settlement curve of a static pile load test either did not exhibit or exhibited a much less obvious second section Normally, in the first section the settlement of a rapid load pile test was larger than that of a static pile load test Figure 6.21 shows that the first section of the rapid load pile test was complete at a settlement of about 0.76mm whereas that of the static pile load test was complete at a settlement of 0.6mm Due to this it is difficult to derive the pile shaft static load-settlement curve from that of a dynamic pile load test since the shape of the load-settlement curve of a dynamic pile load test is not similar to that of a static pile load test

♦ The damping load not only depends upon the shearing rate but also upon the soil loading stages described above It can be seen from Figure 6.10 to 6.24 that the damping load develops gradually from the first section to the final section

♦ Comparing the derived load-settlement curves of the Randolph and Deeks and Gibson and Coyle models it could be said that Randolph and Deeks’ model is only another form of the Gibson and Coyle model

♦ In general, the three models predict damping load well when the second section of the load-settlement curve of a dynamic pile load test develops over a small settlement (Figure 6.13 to 6.15) whereas they overpredict the damping loads of the first and the second sections of the load-settlement curves when the second section develops over a large settlement (Figure 6.19 to 6.24)

♦ Balderas-Meca’s model takes into consideration the development of the damping load between the soil stages by changing the damping parameter, α, with the pile settlement However, quake values for the pile shaft load are over a relatively wide range so that the quality of the derived load-settlement curve is not consistent (Figure 6.12 and 6.15)

♦ The viscous damping that occurs during a rapid load pile test can be represented by a non-linear power law incorporating damping coefficients

♦ Apart from the above models other models reported in the literature review have also been examined However, they did not provide a good prediction for statnamic load tests The main reason is that these models were only proposed for the ultimate shear resistance

♦ The soil damping characteristics depend on soil properties, i.e liquid and plastic limits Summarizing data from previous studies, Hyde et al (2000) showed that the damping parameter, α, can vary by orders of magnitude for different clays

6.3.2 A new non-linear model (the Proportional Exponent Model) for pile shaft rate effects

In pile load tests settlement criterion, which is normally chosen as 10% of the diameter of the pile, rather than the pile ultimate load is used for the determination of a pile’s capacity Therefore, it is desirable to derive the load-settlement relationship for a static condition from that of a statnamic test The most widely used Unloading Point Method (UPM), which assumes that the damping load is linear with the shearing velocity and can overestimate the ultimate static pile capacity by up to 30% for piles in clay The non-linear power laws reviewed in Section 6.3.1 could predict the ultimate static pile bearing capacity well However, in order to get a better estimate the pile’s bearing capacity at loads below the ultimate static pile bearing capacity,

which is the zone of most interest in determining the pile’s serviceability, the available non-linear power laws need to be modified

The available non-linear power laws should be modified in such a way that they can simulate the gradual development of the damping load from the first to the final stages of loading This can be achieved by gradually increasing the damping parameters with the development of the dynamic pile load, and the damping parameters are constant when the pile’s dynamic load reaches the ultimate value Thus, the new non-linear power law should be in the same form as the available models when the pile’s dynamic bearing capacity reaches its ultimate value It is proposed therefore to use a model with a proportional exponent of the velocity term, the general form of which is as follows:

vv τ

(6.4) where τd is the dynamic shear resistance

τs is the assumed static shear resistance determined at a pile velocity of 0.01mm/s

τd(ultimate) is the ultimate dynamic shear resistance

vd is the pile velocity which can vary from zero to 2500mm/s during a statnamic test (Japanese Geotechnical Society, 2000)

vs is the assumed static pile velocity which is 0.01mm/s in this study α, β are the damping coefficients

The damping parameters depend on the soil properties However, from previous researches (see Section 2.9), combined with test results from this study, the damping parameter β = 0.2 can be used for clay Thus Equation 6.4 becomes:

(6.5) The validity of this equation will be examined by using the results of the calibration chamber pile load tests

When the pile shaft resistance reaches the ultimate value τd/τd(ultimate) = 1 and Equation 6.5 becomes:

(6.6) where τd(ultimate) is the ultimate dynamic shear resistance

τs(ultimate) is the ultimate static shear resistance

vd(ultimate) is the shear velocity corresponding to the ultimate dynamic shear resistance

Due to the rate effects of the five clay beds being slightly different as mentioned in Section 6.3.1 the damping parameter will be deduced independently for each bed

Equation 6.6 is used to calculate the damping parameter, α, from the pile shaft load ratio Plots of τd/τs - 1 against vd are shown for Beds 2 to 5 in Figures 6.25 to 6.28 The best match between the test data and Equation 6.6 was achieved using least square regression gives the α values of 0.085, 0.135, 0.115, and 0.145 with the fitting regression coefficient R-squared values of 0.90, 0.88, 0.87, and 0.89 for Bed 2 to Bed 5 respectively The average damping parameter (αaverage = 0.12) is deduced for all five clay beds as shown in Figure 6.29 It can be seen that the use of a damping parameter varying from α = 0.068 to α = 0.16 in Equation 6.6 covers for all five beds’ pile shaft dynamic loads (Figure 6.29) Figures 6.25 to 6.28 plus 6.29 show that damping effects are fairly consistent for each clay bed, but vary from bed to bed This is thought to be attributed to the variation of the materials used as mentioned in Section 5.2

To examine the sensitivity of the damping parameter β, several values of β (0.16; 0.18; 0.22; 0.24) are plotted in Figures 6.25 to 6.28 In general β = 0.18 and β = 0.22 are lower and upper boundaries for the test data

In the following sections Equation 6.5 will be used to derive the full static settlement curves from the rapid load pile tests for Beds 2 to 5 using the damping values above

load-♦ Clay Bed 1: As mentioned in Chapter 3 (see Sections 3.4.6 and 3.4.7) the pile was installed into the clay bed after 14 days of triaxial consolidation under a pressure of 280kPa After pile installation the clay bed continued to be subjected to a pressure of

280kPa This installation during the triaxial consolidation caused disturbance between the pile shaft and the bored hole which could not be eliminated during the subsequent the second stage of consolidation It was found that the pile bearing capacity of this bed was much lower than that of the other beds even though the clay beds underwent the same consolidation history Only the results of one CRP test at a rate of 0.01mm/s (B1/1/CRP-0.01) will be used together with two rapid load pile tests, a CRP test at a rate of 100mm/s (B1/2/CRP-100) and a statnamic pile load test (B1/4/STN-15) since the others tests were mainly carried out to choose the best input parameters for the testing equipment for the pile load tests of the following beds The CRP test at a rate of 0.01mm/s provides the static pile bearing capacity benchmark for the two rapid load pile load tests The derived static load-settlement curves which are obtained from the rapid load pile tests by the proportional exponent model are shown in Figures 6.30 and 6.31 The damping parameter α = 0.14 was obtained for clay Bed 1 from the best fit between the pile’s measured static shaft capacity and the model In terms of the pile’s ultimate bearing capacity Test B1/4/STN-15 had the largest deviation (3.6%) between the measured pile static capacity and the derived pile static bearing capacity (Figure 6.30)

♦ Clay Bed 2: The static load-settlement curves derived from rapid load pile tests by the proportional exponent model are shown in Figure 6.32 to 6.41 The damping parameter α = 0.085 was used for this clay bed In terms of the pile’s ultimate bearing capacity Test B2/19/CRP-150 had the largest deviation (9%) between the measured pile static capacity and the derived pile static capacity (Figure 6.41)

♦ Clay Bed 3: The results of the analysis for pile load tests on Bed 3 are shown in Figures 6.42 to 6.53 The damping parameter, α, derived for this bed was 0.135 As mentioned earlier, the pile’s static bearing capacities obtained from CRP tests at a rate of 0.01mm/s were not constant during testing due to local consolidation following each test For that reason slow rate - CRP tests were repeated after each group of rapid load tests for a given bed However, in some cases there was a large difference in the pile’s static bearing capacity between two adjacent CRP tests at a rate of 0.01mm/s In these cases both CRP tests at a rate of 0.01mm/s are used (Figures 6.42 to 6.44 and 6.46 to 6.49) One CRP test was carried out before the rapid load test,

giving measured static load 1, and the other, giving measured static load 2, after the rapid load pile test In terms of the pile’s ultimate bearing capacity Test B3/16/CRP-300 had the largest deviation (12.9%) between the measured pile static capacity and the derived pile static bearing capacity (Figure 6.51)

♦ Clay Bed 4: From Bed 1 to Bed 3 the clay beds underwent a similar consolidation history, i.e 280kPa pressure for both 1-D and isotropic triaxial consolidations Clay Bed 4 had a similar consolidation history to that of the first three beds only for the first series of tests After the first series of tests the clay bed was subjected to a triaxial consolidation pressure of 400kPa Although two series of tests were carried out at different soil consolidation pressures it was found that one value of damping parameter α = 0.115 could be used for this bed The results of clay Bed 4 are shown in Figures 6.54 to 6.66 In terms of the pile’s ultimate bearing capacity Test B4/26/CRP-50 had the largest deviation (10.8%) between the measured pile static capacity and the derived pile static bearing capacity (Figure 6.66) Although several statnamic pile load tests were carried out in Bed 4 under a consolidation pressure of 400kPa their load-settlement curves are not used for analysis since the pile’s dynamic bearing capacity was large in comparison with the applied load system’s capacity, which was 41 kN, and only small settlements occurred in these statnamic pile load tests

♦ Clay Bed 5: This bed’s consolidation history was different from those of the first four beds The 1-D consolidation pressure for this bed was only 240kPa and three series of pile load tests were carried out at three different 3-D triaxial consolidation pressures which were 240kPa, 280kPa, and 340kPa respectively Similar to Bed 4, although three series of tests were at different clay soil consolidation histories only one damping parameter α = 0.145 was used The results of tests carried out in Bed 5 are shown in Figures 6.67 to 6.81 In terms of the pile’s ultimate bearing capacity Test B5/32/CRP-50 had the largest deviation (8.62%) between the measured pile static capacity and the derived pile static bearing capacity (Figure 6.81)

For a comparison between the new model and the Randolph and Deeks model, attention is given to the pile shaft resistance below the ultimate as shown in Figures

6.82 and 6.83 Figure 6.83 shows that at a settlement of 0.45mm the measured static load was 4.2kN but the derived static load from the Randolph & Deeks model was 3.3kN an underprediction of about 22% whereas the derived static load from the proportional exponent model was 3.98kN and deviated from the measured value only by 5% Pile velocities are also plotted in these figures It can be seen that during both statnamic and CRP tests pile velocities increased gradually with the development of the applied load

6.3.3 Pile shaft softening effect

It has been shown by previous researchers that pile shaft load in clay is fully mobilised at a certain pile settlement and then decreases to a residual value with further settlement The residual resistance was reported in some cases to be as low as 50% of the maximum resistance (Chandler & Martins, 1982) However, for pile design purposes it is recommended that 70% of the maximum pile shaft resistance should be used for the residual pile shaft resistance (API, 1993) This behaviour is attributed to the parallel alignment of the clay particles to the shear direction after a significant deformation (Lemos & Vaughan, 2004), breaking of the clay particle interlocking and breaking the initial cementation between the pile and clay (Mitchell, 1976)

In this study it was found that softening effects were only exhibited in constant rate of penetration tests with the rates varying from 0.01mm/s to 100mm/s The results of tests in which softening occurred have been placed into two groups: i) CRP tests at the rate of 0.01mm/s; ii) CRP tests at rates larger than 0.01mm/s The results of these tests are shown in Figure 6.84 to 6.93

It was found that a relatively large pile settlement was needed for the pile shaft load to reach the residual value However, studying softening effects was not the main object of this study and if the residual load had been reached for every test the numbers of tests for each bed would have reduced significantly Therefore, in several tests the residual load were not reached as the pile settlements were normally controlled to about 7mm

In general, the results show that the degree of softening, defined as the residual shaft load relative to the maximum shaft load, decreased from test to test and in some cases softening effects vanished after several tests (Figure 6.85; 6.86; 6.87; 6.90; 6.92) This suggests that the alignment of the clay particles parallel to the shearing direction played an important role in softening and further softening did not occur after several tests as a result of the particle arrangement reaching a stable condition In addition, softening effects were not found in the pile load tests with shearing rates higher than 100mm/s It is suggested that there is a critical shearing velocity for soil and the modes of shearing are different depending on whether the shearing velocity is higher or lower than the critical shearing velocity If the shearing velocity is higher than the critical velocity the shearing mode is turbulent as defined by Lemos & Vaughan (2000) It suggests that the soil particles do not have enough time to rearrange, and as a result softening effects do not occur On the other hand if the shearing velocity is much lower than the critical velocity the soil particles will have enough time to rearrange and softening will occur If the shearing velocity is just lower than the critical shearing velocity a transitional shearing mode may be exhibited in which the re-arrangement of the soil particles can only develop to a certain degree and in the transitional range the higher the shearing velocity the lower the degree of softening

These results show that the degree of softening was not consistent from bed to bed These effects were exhibited markedly for Bed 3 (Figures 6.88 and 6.88) whereas they were exhibited slightly for Bed 4 (Figure 6.90 and 6.91) This is thought to be attributed to the inconsistency in the use of materials as mentioned in Section 5.2

As mentioned above, the alignment of the clay particles parallel to the shearing direction causes softening It is expected to occur in the first test in any bed and as a result softening effects are likely to decrease from test to test as further tests are carried out in the bed However, softening did not occur for the first test in Beds 3 and 5 (Figures 6.88 and 6.92) but occurred in later pile load tests These results seem to contradict the above-mentioned reason for softening There are no clear explanations for this

Consistent with previous studies (Chandler & Martins, 1982), the pile shaft residual load could be as low as 67% of the maximum pile shaft load (Pile load test B3/2/CRP-

10 in Figure 6.89) In practice, it is appropriate and conservative if softening is taken into consideration to deduce the maximum pile load by 30% as recommended by API

In this study rate effects were not separated from softening effects at ultimate condition However, the influence of the softening effects is not significant compared to rate effects due to the following reasons:

♦ Softening effects did not occur during rapid load tests with the shear rates above 100mm/s

♦ The softening effect only occurred when the pile resistance reached the ultimate value Therefore, below the ultimate value rate effects were not influenced by softening effects

♦ The residual resistances in CRP tests at the rate of 0.01mm/s, which were considered as the static benchmark, were only different from the ultimate values by about 5%

6.3.4 Repeatability of the static pile shaft loads

The measured pile shaft loads for static pile load tests are shown in Table 6.1 In this table some pile load tests have two values for the shaft load The first value is the maximum pile shaft load and the second value is the residual pile shaft load as a results of softening It can be seen from the table that the pile shaft loads of the first test of Beds 2 to 4 were fairly consistent They only varied from 3.50kN to 3.71kN With Bed 5, due to the clay bed being subjected to 1-D and isotropic triaxial consolidation pressures of 240kPa instead of 280kPa the pile shaft load for the first pile load test was only 2.90kN The second CRP test at a rate of 0.01mm/s in each bed was carried out after several rapid load tests and as a result of the local consolidation, which developed around the pile during these tests, the pile shaft loads increased considerably The pile shaft loads of the second CRP tests at the rate of 0.01mm/s in Bed 2, 3, and 4 were 4.41kN, 5.10kN, and 4.58kN If the average value of the pile shaft loads, 4.69kN, is used as a benchmark the maximum deviation from the benchmark is 8.4% The pile shaft loads of pile load tests at the rate of 0.01mm/s in Bed 5 under a triaxial consolidation pressure of 280kPa were 4.32kN to 4.42kN and they were quite consistent with those of later tests in Bed 2 to 4 Normally, the

maximum pile shaft loads of CRP tests at the rate of 0.01mm/s in each clay bed were achieved for the final pile load tests The maximum pile shaft loads in Beds 2 to 4 under the isotropic triaxial consolidation pressure of 280kPa were 5.41kN, 5.10kN, 4.81kN If the average value, 5.11kN, is used as a benchmark the maximum deviation from the average value is 6%

Four isotropic triaxial consolidation pressures (240kPa, 280kPa, 340kPa, and 400kPa) were used for the five clay beds It was found that the ultimate pile shaft load was in proportion to the radial effective stress The ratios of the ultimate unit shaft load to the radial effective stress of the first test in Beds 2 to 5 are 0.21, 0.21, 0.21, and 0.20 respectively (Table 6.1) In general these ratios increased for the following pile load tests due to the development for the local consolidation of the soil around the pile shaft and the maximum ratios of Beds 2 to 5 were 0.30, 0.30, 0.29, and 0.27 respectively

Comparing the ratio of the maximum pile shaft loads to the radial effective stresses and to the undrained soil shear strengths which are shown in Tables 5.4 to 5.7 it is found that the relationship between the maximum pile shaft load and the radial effective stress is more consistent than that between the maximum pile shaft load and the undrained soil shear strength It suggests that the effective stress design method proposed by Chandler (1968) and Burland (1973) to determine the maximum pile shaft load for clay seems to be better than the total stress design method proposed by Skempton (1959)

6.4 Pile tip load results

Normally for clays the pile tip bearing capacity is small in comparison with the pile’s shaft bearing capacity Therefore, when deriving a pile’s static bearing capacity from a rapid load pile test attention is mainly given to the pile’s shaft load The pile tip load was measured by a pile tip load cell (see Section 3.5.1) Pile tip load results for Bed 1 to Bed 4 are shown in Tables 6.2 to 6.5 respectively The pile tip load cell did not work properly for Bed 5 so its pile tip load results are not reported

In several early tests in each bed the pile tip load was fully mobilised at a quake of about 10%-12% of the pile diameter (70mm) (Figure 6.94) Due to a residual tip load locked into the soil below the pile base this quake dropped gradually after the first few tests and finally it was as low as 2.5%-4% of the pile diameter (Figure 6.94)

A pile tip residual load occurred before the first to the final pile load tests It is likely that the residual pile tip load occurred before the first test of each bed due to the fact that under the isotropic triaxial consolidation pressures an upward force occurred at the pile tip In the following tests, a residual pile tip load was generated for the following reason additional to the above During the test, the load cell was compressed and the pile tip load governed the elastic deformation of the load cell (Figure 6.95) When the test had finished, the load cell had a tendency to return to the unload condition This means that it needed to expand to its normal length However, when it expanded it was resisted at one end by soil at the pile tip and at the other end by the upper part of the pile; in turn the upper part of the pile would have mobilised shaft friction to resist this effect Because of this, the load cell did not expand totally and the pile tip residual load depended on the soil properties at the pile tip and pile shaft friction So when the pile tip residual loads were recorded the soil at the pile tip was compressed with a load equal to the residual load and the pile shaft friction had a value equal to the pile tip residual load minus pile weight However, the shaft friction load cell was not designed to measure this value By and large, the pile working with a residual tip load had a mechanism similar to that of a pile with an Osterberg cell test at its base For this reason, the soil beneath the tip of the pile was under two different loads during a test The first load was a static load equal to the residual load and the second was the subsequent testing load

The residual pile tip load seemed to increase over the first few tests and then it became stable (Tables 6.2 to 6.5) This can be explained from the mechanism which generated the residual tip load The pile and soil compressions at the pile tip during a pile load test are locked by the pile shaft load when the test finishes Thus, the magnitude of the residual pile tip load depends on both soil properties around the pile shaft and at the pile tip For the first few pile load tests the soil around the pile shaft and at the pile tip developed local consolidation so that residual tip loads increased and became stable when this local consolidation was complete

The pile tip load in static pile load tests, CRP tests at a rate of 0.01mm/s and maintained load tests, can be summarized as follows:

♦ Bed 1: It increased from about 3.69kN (B1/1/CRP-0.01) to about 5.5kN (B1/12/CRP-0.01) Due to the soil in the vicinity of the pile base being very stiff the hand vane test could not determine its shear strength when the clay bed was dissected The soil shear strength at the pile base can be determined indirectly if the following pile base bearing capacity equation for clay is used (Meyerhof, 1952):

Qb = cub Nb Ab (6.7) where Qb is the pile tip bearing capacity

cub is the undrained shear strength of the soil at the vicinity of the pile tip Nb is the pile tip bearing capacity factor which is 9 for a deep circular footing clay

Ab is the pile tip cross sectional area

Using Equation 6.7 the undrained shear strength of Bed 1 varied from 107kPa to 159kPa

♦ Bed 2: The pile tip load varied from 3.79kN (B2/1/CRP-0.01) to 7.27kN (B2/16/CRP-0.01) It was noted that after three consecutive maintained pile load tests local consolidation of the soil at the pile tip developed significantly as the pile tip load increased from 6.57kN (B2/12/CRP-0.01) before the maintained load tests up to 7.27kN (B2/16/CRP-0.01) after the maintained load tests Applying Equation 6.7 the undrained shear strength of the soil at the pile base varied from 109kPa to 210kPa

♦ Bed 3: The pile tip load varied from 3.73kN (B3/1/CRP-0.01) to 6.82kN (B3/22/MLT) giving a variation of the undrained shear strength of the soil at the pile tip from 108kPa to 197kPa using Equation 6.7

♦ Bed 4: There were two series of pile load tests for Bed 4 The first series was carried out when the 3-D consolidation pressure of the clay bed was 280kPa The pile tip loads varied from 4.09kN (B4/1/CRP-0.01) to 6.59kN (B4/17/CRP-0.01) and using Equation 6.7 the undrained shear strength of the soil at the pile tip varied from 118kPa to 190kPa The second series was carried out when the clay bed had a 3-D

consolidation pressure of 400kPa The pile tip loads varied from 8kN to 8.5kN and using Equation 6.7 the undrained shear strength of the soil at the pile tip varied from 231kN to 245.5kN

It can be seen that in terms of the pile tip static bearing capacity a high consistency was achieved for the pile tip loads of the first test in each beds The pile tip loads were 3.69kN, 3.79kN, 3.73kN, 4.09kN for Bed 1 to Bed 4 respectively

The undrained shear strengths of the soil at the pile base obtained using Equation 6.7 seem reasonable when compared with the soil shear strengths which were obtained in the vicinity of the pile shaft in Bed 4 by hand vane tests (Table 5.6) The shear strength of the soil at the pile shaft could be as large as 108kPa and the shear strength of the soil at the pile tip was found higher than that

The pile tip damping loads, defined as the difference between the pile tip load of a static load test and a rapid load pile test are shown in Tables 6.2 to 6.5 The pile tip static loads between two CRP tests at the rate of 0.01mm/s are interpolated linearly with the number of tests from the measured pile tip loads in these two tests Similarly, the pile tip damping load seems to be non-linear with the pile velocity in contrast to the linear pile base model proposed by Randolph and Deeks (1992) which can be expressed in the form of:

Fb = b + b + b (6.8) where Fb is the total base load

Kb is the pile base’s spring stiffness which is given by Equation 6.9 w is the pile tip settlement

Cb is the dashpot constant which is given by Equation 6.10

Mb is the lumped mass soil at the pile tip which is considered as a part of a pile tip and is given by Equation 6.11

a is the pile tip acceleration

(6.10)

Mb (6.11) where G is the soil shear modulus

D is the pile diameter μ is the Poisson’s ratio ρ is the soil bulk density

However, no new equation is proposed to quantify the pile tip damping load for the following reasons:

♦ The development of the pile tip load during a rapid load pile test was not a totally dynamic process since a relatively high residual load always existed Therefore, it could be considered that the soil at the pile tip underwent two steps during a rapid load test, a static load due to the residual load and then a subsequent dynamic loading process

♦ The pile tip load cell was subjected to the residual load for a long period of time in each clay bed Normally it was about 2 months from the pile installation until the clay bed was stripped down Due to this, creep occurred to the pile tip load cell and it altered the zero load reading of the pile tip load cell This was recognized when there was a difference of the zero load reading of the load cell before the pile installation and after the bed had been stripped down This problem became more severe after the maintained load tests The alteration of the zero load reading made the pile tip loads unreliable for the quantification of tip damping load

6.5 Application of the proportional exponent model to the pile total load

As mentioned in Section 6.4 the measured pile tip loads were not reliable Therefore, no attempt has been made to build a model for the pile tip load However, in practice the model for the pile tip bearing capacity should not be too important for a pile installed in clay as the pile tip bearing load is normally much smaller than the pile

shaft bearing capacity As a simplification the model for the pile shaft load can be applied to the total pile load

This section will use the new model which was proposed in Section 6.3.2 and expressed by Equation 6.5 to derive the pile static bearing capacity from the pile dynamic bearing capacity The inertial forces of the pile were taken into consideration Although in practice it was found that they were negligible as the model pile weight was only 19kg The damping parameters, α, for Beds 1 to 5 are those which were used for the pile shaft loads and the results are shown in Figure 6.96 to 6.100 It can be seen from these figures that the model can be applied fairly well to find the total static pile load-settlement curve

Brown (2004) laboratory test data was also used to check the model Due to the same test method having been applied for the two studies an average damping parameter, α = 0.12, was used to derive the equivalent static load-settlement curve (Figure 6.101) It can be seen from the figure that the model works well for Brown (2004) data and this gives a confidence in the consistency of the test method

6.6 A simple theoretical approach for the load transfer mechanism

In this section a simple theoretical method will be developed which can be employed to establish the relationship between the pile settlement and its shaft load The method was suggested by Seed and Reese (1957) and then developed by Randolph and Wroth (1978); Kraft et al (1981a) and Armaleh and Desai (1987)

6.6.1 Available models for load transfer

The method was developed from the following assumptions:

♦ The pile shaft will transfer shear stresses to the soil and to a certain distance from the pile called the influence zone beyond which these shear stresses are negligible ♦ The load-settlement behavior of the pile shaft may be considered separately from that of the pile base The interaction between the soil above and below the pile base level is taken into account by reducing the influence zone, rm, to a limited radius (Figure 6.102)

♦ The displacement pattern of the soil around the pile shaft can be considered as concentric cylinders in shear (Figure 6.102) Radial soil displacements due to pile loads are assumed negligible when compared to vertical soil deformations Thus, a simple shear condition prevails in the soil

♦ The influence zone, rm, is given by the following empirical formula (Randolph and Wroth, 1978):

rm (6.12) where L is the length of pile shaft embedded in the soil

ρ is the heterogeneous factor defined as the ratio of the shear modulus of the soil at the depth of L/2 to that at the depth of L

μ is the Poisson’s ratio

Consider a soil element which is a radial distance r away from the pile centre (Figure 6.102) The vertical deformation of the element, dw, is given by:

γ = = ⇒ = (6.13) where γ is the shear strain

G is the soil’s shear modulus

τr is the shear stress on the soil element which is a radial distance of r away from the pile centre

The vertical settlement at the pile shaft, wro, can be obtained by integration of the element deformation given by Equation 6.14 from the pile shaft with a radius of ro to the influence zone, rm

= m

(6.14) The shear stress, τr, on the soil element which is away from the pile shaft a distance of r can be obtained from the equilibrium condition (Figure 6.102) and it is given by:

τ = (6.15) where τrois the shear stress at the pile shaft and ro is the pile radius

Substituting τr from Equation 6.15 to Equation 6.14 the vertical deformation at the pile shaft is given by:

(6.16) If the soil shear modulus, G, is considered constant then Equation 6.16 gives a linear relationship between the pile deformation and the pile shaft load In reality the relationship is non-linear To take this non-linearity into consideration Kraft et al (1981a) developed the model by proposing a hyperbolic expression for the soil’s shear modulus:

G= − (6.17) where G is the shear modulus at an applied shear stress τo; Gi is the initial shear modulus at small strains (Gi = Ei/[2(1+μ)])

Rf is the stress-strain curve-fitting constant which should be taken from 0.9 to 1 (API, 1993)

τmax is the shear stress at failure μ is the Poisson’s ratio

Using the new model proposed by Kraft et al (1981a) the vertical deflection at the pile shaft is given by:

o (6.18)

Equations 6.16 and 6.18 give a relationship between the pile shaft load and the pile settlement for a static pile load test In the following section some modifications will be suggested to the model for static tests and then the model will be developed for rapid load pile tests

6.6.2 Modifications to the existing models for load transfer for static pile load tests and a new model for rapid load pile tests

To apply this method to the pile load tests of this study a modification is proposed to the influence zone, rm, which was originally suggested by Randolph and Wroth (1978), and then the method is developed for rapid load pile tests It should be made clear that the method establishes the relationship between the pile shaft load and the

pile shaft deflection before failure and no attempt will be made to take the post failure softening effect into consideration

Randolph and Wroth (1978) proposed the influence zone had a radius which was given by Equation 6.12 and that it remained constant during the development of the pile shaft resistance Other researchers still use this assumption However, in reality it is believed that the radius of the influence zone, rm, depends on the pile shaft resistance and increases with the development of the pile shaft resistance The influence zone reaches a maximum value when the pile shaft resistance reaches the ultimate pile shaft resistance For simplicity of the subsequent integrations a simple influence zone equation is proposed

rm = rormmax +ro

(6.19) where rm is the influence zone when the pile shaft resistance is τro

τmax is the ultimate pile shaft resistance

rmmax + ro is the maximum influence zone when the pile shaft resistance reaches τmax

If it is assumed that the soil’s shear modulus is constant during loading the vertical settlement at the pile shaft of a static pile load test is given by:

(6.20)

If it is assumed that the soil’s shear modulus follows the hyperbolic model proposed by Kraft et al (1981a) then the pile shaft settlement of a static load pile test is given by:

(6.22)

However, the distribution of the shear stress away from the pile shaft is different from that of a static pile load test, which is given by Equation 6.15, if the inertial force is taken into consideration The equilibrium condition for a soil element (Figure 6.102) is then given as follows:

rrdrSBBCCr ⎟ ' '

⎝⎛ ++

where S is the symbol denoted for areas (Figure 6.102) σz is the vertical stress

γsoil is the soil bulk density

ar is the acceleration of the soil element which is a radial distance of r away from the pile centre

V is the volume of the soil element m is the mass of the soil element

The areas, volume, and mass in the Equation 6.23 are given by the following equations:

⎝⎛ +=

SABCDBCD θ (6.24)

SAA'D'D = θ (6.25)

SBB'C'C =( + ) θ (6.26)

⎝⎛ +=

' θ (6.27)

⎠⎞⎜⎜⎝⎛= γ

(6.28) Substituting the areas, volume, and the soil mass of the element given by the above equations into Equation 6.23 the equilibrium condition for the element becomes:

(6.29) Ignoring the vertical stress which is induced by the pile shear stress, the following expression can be obtained:

dσ =γ (6.30)

Substituting Equation 6.30 into Equation 6.29 the equilibrium condition becomes:

(6.31) The acceleration of the soil element can be obtained from the pile acceleration, aro, by using Equation 6.32 with the assumption that the soil deformation at a location in the influence zone is reciprocally proportional to its distance from the pile shaft

τ (6.34) where A, B, and C are parameters which can be obtained by substituting Equation 6.34 into Equation 6.33 and these parameters are given as follows:

A=τ − − ⎜⎜⎝⎛γ ⎟⎟⎠⎞ + − ⎜⎜⎝⎛γ ⎟⎟⎠⎞)

(6.35)

(6.36)

C =− − ⎜⎜⎝⎛γ ⎟⎟⎠⎞)

2 (6.37) Substituting A, B, and C given by Equations 6.35 to 6.37 into Equation 6.34 the distribution of shear stress can be obtained as follows:

arg

)(

If the soil shear modulus, G, is considered constant, and using the shear stress on the element soil at a radial distance of r from the pile centre given by Equation 6.38, the pile shaft settlement in a rapid load pile test can be obtained

= m

(6.39) hence

= m

w 2 1 (6.40) or

= (6.41)

or

(6.42) However, comparing the pile shaft settlements with and without the inertial force being taken into consideration the influence of the inertial force is negligible Using different soil properties and different pile accelerations for two models the difference between them is less than 5% (Table 6.6) For this reason and for simplicity the following analysis for a rapid load pile test will ignore the influence of the inertial force and Equation 6.15 will be taken for the equilibrium condition

Observing the load-settlement curves of a static and a rapid load pile test (Figure 6.59) it is noticed that soil’s shear modulus is higher for a rapid load pile test than for a static pile load test since under the same load the pile settlement in the static load test is larger than that in the rapid load pile test Considering a linear relationship between shear stress and shear strain the following model is proposed for the rapid load pile test

(6.43)

If the non-linearity between shear stress and shear strain is taken into consideration then the model can be expressed in the form of:

is the ultimate dynamic shear stress

α, β are the damping parameters and β can be taken as 0.2 for clay

For simplicity in obtaining the pile shaft settlement it is reasonable to assume that the shear velocities of soil elements around the pile shaft, vr, equal the shear velocity of a soil element at the pile shaft, vro In addition, the shear stresses of soil elements around the pile, τr, are assumed to equal the shear stress of a soil element at the pile shaft, τro, in the power component, β(τr/τmax

), of Equation 6.43 and 6.44 Thus, Equation 6.43 becomes Equation 6.45 and Equation 6.44 becomes Equation 6.46

(6.45)

(6.46)

Using Equation 6.45 and 0.2 for the damping parameter, β, the pile shaft settlement in a rapid load pile test assuming a linear shear stress-shear strain relationship is given as follows:

(6.47)

(6.48)

τ (6.49)

Similarly, using Equation 6.46 and 0.2 for the damping parameter, β, the pile shaft settlement of a pile rapid load pile test assuming a non-linear shear stress-shear strain relationship is given as follows:

⎟⎟⎠⎞⎜⎜⎝⎛+= m

⎟⎟⎠⎞⎜⎜⎝⎛+= m

τ (6.51)

o (6.52)

oo

In summary two equations are suggested for static tests and two equations for rapid load tests Equation 6.20 presents a linear model and Equation 6.21 presents a non-linear model for static pile load tests Equation 6.49 presents a linear model and Equation 6.53 presents a non-linear model for rapid load pile tests In the following sections these models will be used for the pile load tests in this study

6.6.3 Application of the models to static pile load tests

In this section the pile shaft load-settlement curves of static pile load tests will be derived based on the model that has been developed in Section 6.6.2 The pile shaft settlement can be obtained from the pile shaft resistance by Equation 6.54 if a linear relationship is assumed between stress and strain or by Equation 6.55 if a non-linear relationship is assumed between stress and strain

(6.54)

(6.55)

In reality the soil shear modulus varies along the pile, with radial distance from the pile and the soil shear modulus variations depend on pile installation methods In this study the variation of the soil shear modulus along the pile was negligible since the pile shaft load was only measured over a length of 302mm After dissecting the clay beds it was recognized that the soil shear strength decreased away from the pile over a length of about two to three pile radii and then became constant However, for simplicity only an average soil shear modulus will be used No experiment was carried out to determine the soil shear modulus for this study However, it was found that the ratio of shear modulus to maximum shaft friction, G/τs, of about 250 can give a good match between the models and the experimental data

Randolph and Wroth (1978) proposed the influence zone given by Equation 6.12 This study only worked with a small size instrumented model pile installed in small scale clay beds which were prepared in the calibration chamber Therefore, no attempt was made to establish an influence zone which can be used universally However, it seems that the influence zone given by Randolph and Wroth (1978) still lacks an element which links it to the soil strength In other words it is believed that the higher the soil strength the larger will be the influence zone Here the influence zone varied from 3 to 5 pile radii for static pile load tests depending on the ultimate pile shaft resistance (Equation 6.19) These values were deduced from the assumption that the influence zone was slightly larger than the local overconsolidated zone around the pile (2 to 3 pile radii) which was determined by moisture content and hand vane tests when stripping down the beds The influence zone was slightly larger than the local overconsolidated zone around the pile as the stresses induced by the pile shaft resistance at the outer zone of the influence zone could not cause significant overconsolidation of the soil

Applying both the linear model, Equation 6.54, and the non-linear model, Equation 6.55, with parameter Rf = 0.99 (API, 1993), to CRP tests at a rate of 0.01mm/s the results are shown in Figures 6.103 and 6.104 The results show that the non-linear model for the relationship between resistance and settlement of the pile is better than the linear model for the first static pile load test in each bed On the other hand, from the second static pile load test in each bed the linear model works better than the non-linear model This could be due to the fact that the soil’s non-linear characteristics decreased when the local consolidation of the soil around the pile shaft developed The failure occurred with less settlement when soil’s stiffness increased

6.6.4 Application of the models to rapid load pile tests

As in Section 6.6.3, the pile shaft load-settlement curves for the rapid load pile tests have been derived by using the linear and non-linear models given by Equations 6.56 and 6.57 The parameter, α, is obtained from Section 6.3, and 0.99 is used for parameter Rf Due to the pile shaft resistances of the rapid load pile tests being higher

than those of the static, the influence zone for rapid pile tests is chosen from about 6 to 9 pile radii depending on the ultimate shaft resistance

τ (6.56)

The results obtained using these equations are shown in Figures 6.105 to 6.107 Equation 6.57 always gives a better relationship between the shaft load and the settlement of the pile shaft than Equation 6.56 This maybe due to the influence zones of the rapid pile tests being larger than the overconsolidation zones around the piles

In summary, it is suggested that:

♦ The non-linear model is more suitable for normal consolidated clay than the linear model

♦ The linear model is more suitable for overconsolidated clay than the non-linear model

In the experiment performed:

♦ The influence zone in the static load test was of similar size to the overconsolidated zone around the pile Therefore, the linear model is more suitable for the static load tests than the non-linear model

♦ The influence zone in the rapid load tests was much larger than the overconsolidated zone around the pile, therefore encompassing a large zone of normally consolidated clay Therefore, the non-linear model is more suitable for the rapid load tests than the linear model

6.6.5 Quake value for the pile shaft resistance of a rapid load test

The settlement at which the pile shaft load is fully mobilised is called the quake value Theoretical (Randolph and Deeks, 1992) and field studies (Tomlinson, 2001) show that the quake values for the pile shaft load of a static pile load test vary from 0.5%-1% of the pile diameter This section will consider the quake values for the pile shaft load of a rapid load pile test

Assuming the non-linear model given by Equation 6.57 can demonstrate the relationship between the pile shaft load and the pile shaft settlement of a rapid load pile test, when the pile shaft load reaches the ultimate soil shear strength the pile shaft deflection reaches the quake, qs, and Equation 6.57 becomes:

ln1 α

(6.58)

In order to compare the quake value of a static pile load test with that of a rapid load pile test the ultimate soil shear loads of a rapid and a static pile load test are denoted by τmaxd and τmax

s respectively Similarly the influence zones of a rapid and a static pile load test are denoted by rmaxm(d) and rmaxm(s) respectively Thus, the quake values of a rapid load pile test, qs(d), and of a static pile load test, qs(s), can be given by Equation 6.59 and Equation 6.60 respectively

(6.60) From Equation 6.59 and 6.60 the relationship between quakes of a static and a rapid load pile test can be expressed in the form of:

(6.61)

If it is assumed that the variation of the radius of the influence zone is linear with the ultimate pile shaft resistance the following expression can be obtained:

(6.62) Substituting Equation 6.62 into Equation 6.61 the following is obtained:

(6.63)

Substituting (rmaxm(s)/ro = 3 and 5), (vd = 100mm/s and 300mm/s), (Rf = 0.99) and (α = 0.085, 0.145) into Equation 6.63 it gives qs(d)/qs(s) = 1.08 and 1.12 respectively It can be recognized that quake of a rapid load test is not much different from that of a static pile load test if the non-linear model is applied to them However, as mentioned in Section 6.6.3, the linear model works better for static pile load tests due to the local consolidation of the soil around the pile shaft, whereas the non-linear model is better for rapid load tests due to the influence zones in rapid load tests being much larger than the overconsolidation zone around the pile If so, applying the non-linear model to rapid load pile tests and the linear model to static pile load tests the ratio of two quake values becomes:

(6.64)

Substituting (rm(s)max/ro = 3 and 5), (vd = 100mm/s and 300mm/s), (Rf = 0.99) and (α = 0.085, 0.145) into Equation 6.64 it gives qs(d)/qs(s) = 3.7 and 4.7 respectively The

ratios vary from about 3.7 to 4.7 matching fairly well with the experimental results (see quakes in Figures 6.104 to 6.106) However, it must be emphasized that Equation 6.64 only applies to the pile load tests of this study due to a particular soil condition which occurred during a series of pile load tests In addition, the calculations of the quake values for the static and rapid load pile tests are based on the assumption that no slippage between the pile and soil occurs before the pile shaft load reaches its maximum value In reality, this slippage could occur in rapid load pile tests and as a result the quake value of rapid load pile tests could be larger than that of the calculations

6.7 A comparison between maintained load tests and CRP tests

In order to compare the CRP tests at the rate of 0.01mm/s with the maintained load tests a CRP test was carried out just before the maintained load tests (Tables 4.2, 4.3, and 4.5) Load-settlement curves for both CRP tests and maintained load tests are plotted in Figures 6.108, 6.110, and 6.111 The pile tip load cell was not reliable for Bed 5 Therefore, the pile tip load was not given in Figure 6.111 The results show that the pile load bearing capacities under maintained load tests were slightly higher than those under CRP tests Figures 6.108 and 6.110 indicate that the local consolidation at the shaft load cell level was complete when the maintained load tests were carried out since no increase of the measured shaft loads was found The local consolidation continued to develop for the soil at the pile tip and for the soil around the pile shaft at the level higher than the pile shaft load cell On the other hand, Figure 6.111 shows that the local consolidation of the soil around the shaft load cell continued to develop during the maintained pile load test and as a result the measured shaft load of the maintained load test was higher than that of the CRP test The pile shaft loads of the maintained pile load tests and CRP tests at a rate of 0.01mm/s also show that rate effects were not exhibited for CRP tests at a rate of 0.01mm/s and the choice of 0.01mm/s, as a static benchmark is appropriate Two consecutive maintained load tests were carried out in Bed 2 and the results are shown in Figure 6.109 The pile bearing capacity of the second test was slightly higher than that in the first test and the increase in the pile bearing capacity was mainly from the pile tip and the pile shaft above the pile shaft load cell

6.8 Pore water pressures around the pile during pile load tests

For each pile load test three phases of pore water pressures will be considered: i) prior to the pile loading; ii) during the pile loading; iii) after the pile unloading The initial locations of transducers in the clay beds are reported in Figures 4.1 to 4.5 and final locations in Figures 5.7 to 5.11

There are four main elements which might cause excess pore water pressure generation around the pile during and after a pile load test

♦ Outward movement of soil to accommodate the pile volume leads to an increase in mean total stress The soil displacement may be considered as a spherical cavity expansion pattern ahead of the pile tip, emerging to a cylindrical cavity expansion along the pile (Randolph, 2003)

♦ Shearing and local remoulding generate changes in mean total stress which will create positive pore water pressures for normally consolidated and lightly overconsolidated clay and negative pore water pressures for potentially dilatant, heavily overconsolidated clay (Randolph, 2003)

♦ The pile shaft load and the pile tip load during the pile loading generate vertical stresses in the soil which will create positive pore water pressures The vertical stress at a given point due to the pile shaft load and the pile tip load may be determined by Mindlin’s solution (Mindlin, 1936) In a homogenous soil this vertical stress increases with the increase of depth when the point under consideration is still above the pile base level, and decreases with the increase of the radial distance from the pile shaft ♦ The pile tip residual load which still exists when a pile load test finishes can cause a long term excess pore pressure at the pile tip

The influence of the above on changes to pore water pressures is different in different soil zones around the pile Therefore, the pore water transducers were placed in four groups:

♦ The pore water pressure transducer at the pile shaft ♦ The pore water pressure transducers around the pile shaft ♦ The pore water pressure transducer at the pile tip

♦ The pore water pressure transducers below the pile tip

6.8.1 Pore water pressures during CRP tests at a rate of 0.01mm/s

To illustrate the pore pressures during CRP tests at a rate of 0.01mm/s two pile load tests are used for each bed, i.e the first CRP test at a rate of 0.01mm/s and a CRP test at a rate of 0.01mm/s after several pile load tests The results are shown in Figures 6.112 to 6.115 On some occasions one or more of the pore water transducers malfunctioned and their results are not shown in the figures

6.8.1.1 Pore water pressures at the pile shaft

The pile shaft pore water pressure transducer was located 534mm above the pile tip to measure pore water pressure at the interface between the pile shaft and the clay The pile only advanced a relatively short distance, about 7mm during a pile load test Therefore, the excess pore water pressure at the pile shaft due to the outward movement of the soil was negligible since there was no outwards movements of soil at the level of this transducer It was likely that the excess pore water pressure monitored by this transducer was governed by shearing and local remoulding and the excess pore water pressures due to the vertical stresses created by the pile shaft load and the pile tip load as the shear stress near the pile surface was significant The pile shaft excess pore pressures and the total pile load are plotted against time in Figure 6.112 From the results the following points can be deduced:

♦ Under shear stresses at the pile shaft the adjacent soil seemed to reduce in volume, or compress at an early stage of loading Therefore, positive excess pore pressures developed at this stage However, when the shear stresses progressively developed the soil at the pile shaft changed from reducing in volume to increasing in volume, or dilating, and as a result positive excess pore pressures dropped and could become negative In general the negative excess pore water pressures in the later tests were larger than those in the first test in each bed This means that the degree of soil dilation depends on the degree of consolidation and the local consolidation around the pile shaft developed from test to test

♦ When the pile load tests finished the shear stresses suddenly dropped to zero and the soil volume seemed to recover some of the volume change that occurred during

shearing, and as a result excess pore water pressures increased when the shear stress was released

♦ Excess pore water pressures were mainly due to shearing and local remoulding and due to the vertical stresses in the soil created by the pile shaft load and the pile tip load and they were transient They developed during the pile load test as a result of the development of the pile shaft load When the pile load test finished they gradually returned to the prior to test values as the pile shaft loads dropped to zero

6.8.1.2 Pore water pressures around the pile shaft

The excess pore water pressures around the pile shaft are plotted along with the total pile load against time in Figure 6.113 From these results the following conclusions can be deduced:

♦ For the first test in each bed mostly positive excess pore pressures occurred during the pile load tests

♦ The positive excess pore pressures of the first tests were larger than those of the later tests although the pile bearing capacities of the later tests were larger than those of the first tests This may be due to progressive consolidation The Skempton pore water pressure coefficient, A, depends on the overconsolidation ratio (OCR) (Barnes, 2000) The excess pore pressure can be calculated by:

Δu = B[Δσ3 + A(Δσ1-Δσ3)] (6.65) where Δu is the excess pore water pressure

B is the pore pressure coefficient which can be considered as 1 for a saturated soil

Δσ3 is the increment of the minor principal stress, σ3 A is the pore pressure coefficient

Δσ1 is the increment of the major principal stress, σ1

Although no clear relationship between OCR and pore pressure coefficient A for stress levels before failure exists, a trend of decrease in A with increase of OCR is exhibited at failure for clays For this reason, it is clear that the higher the overconsolidation ratio the lower the excess pore water pressure under the same applied stress

♦ For the later pile load tests negative excess pore pressures occurred in the soil zones near the pile shaft This shows that the soil near the pile shaft was overconsolidated and dilation occurred during shearing However, these negative excess pore pressures changed to positive when the shear stress increased This may be due to the vertical stresses in the soil created by the pile shaft load and the pile tip load offsetting the negative excess pore pressures due to dilation

6.8.1.3 Pore water pressures at the pile tip

A pore pressure transducer was mounted at the pile tip to measure the pore pressure at the interface between the pile tip and soil The excess pore pressures measured at the pile tip along with the total pile load are plotted against time in Figure 6.114 From these results some points can be deduced:

♦ Only positive excess pore pressures occurred at the pile tip This was due to compression only taking place at the pile tip

♦ At the pile tip the excess pore water pressures of the later pile load tests were much lower than those of the first pile load tests, although the applied loads of the later tests were larger than those of the first tests This may be due to the Skempton pore water pressure coefficient, A, decreasing with the increase of the overconsolidation ratio as mentioned in Section 6.8.1.2

♦ When the pile load test finished a large excess pore water pressure still existed This is mainly due to two reasons: i) A residual tip load existed when the pile load tests finished ii) Outward movement of soil occurred to accommodate the volume of the pile moving into soil Thus, consolidation took place at the pile tip under these excess pore pressures

6.8.1.4 Pore water pressures below the pile tip

From Bed 3 onwards one pore water pressure transducer was positioned in the clay below the pile tip The data from this transducer are shown in Figure 6.115 From these results the following points can be deduced:

♦ The excess pore water pressures below the pile tip were quite similar to those at the pile tip due to the similarity in the stress state

♦ The excess pore water pressure below the pile tip was mainly due to the vertical stress on soil due to the pile shaft and tip loads However, the vertical stress generated by the pile tip load was dominant since this zone was near the pile tip

♦ The influence of the pile tip residual load on the transducer below the pile tip was negligible Therefore, the excess pore pressures at this transducer dropped quickly to zero when the pile load tests were complete This was consistent with the overconsolidated clay zone at the pile tip which was about three pile radii, i.e 105mm This overconsolidated clay zone was determined by visual observation of the soil properties when the clay beds were dissected and by moisture contents at different locations away from the pile tip (see Section 5.5)

♦ At the pile tip the excess pore water pressures in the later pile load tests were smaller than those in the earlier pile load tests even though the pile load of the later pile load tests were larger than those of the earlier pile load tests This is thought to be attributed to the soil local consolidation The local consolidation at the location of the pore water pressure transducer below the pile tip was much less than that at the pile tip and it can be seen that the excess pore pressures were proportional to the pile loads (Figure 6.115)

6.8.2 Pore water pressures during maintained pile load tests

The excess pore water pressures measured during a maintained pile load test are shown in Figure 6.116 In contrast to the CRP tests at a rate of 0.01mm/s the behavior of the excess pore pressures during maintained load pile tests was similar at different locations around the pile All transducers gave the following characteristics:

♦ Positive excess pore water pressures were generated under each loading increment These pore water pressures dissipated gradually under each maintained load

♦ When the pile load test was completed the excess pore water pressures dropped from positive to negative It is likely that these negative excess pore pressures were due to swelling when the pile was unloaded

6.8.3 Pore water pressure regime during rapid load pile tests

Rapid load tests included both CRP pile load tests at high rates of penetration and statnamic pile load tests They were carried out over short periods of time (less than 1 second) Therefore, it should be noted that milliseconds are chosen to present time

It was found that the top cell pressure fluctuated during rapid load pile tests It dropped from 280kPa to 225kPa before the connections between the loading frame and the chamber body were strengthened and from 280kPa to 255kPa after the connections were strengthened (see Section 5.4) However, as demonstrated in Figures 5.13 and 5.14 the influence of the top cell pressure fluctuations on the clay bed pore water pressures were negligible

6.8.3.1 Pore water pressures at the pile shaft

The excess pore pressures of the transducer at the pile shaft for several rapid load pile tests are shown in Figure 6.117

The negative excess pore pressure developed as the shear stress increased This indicated that the soil near the pile shaft dilated during shearing Similar to the static pile load tests the changes in the pore water pressures reduced when the local consolidation around the pile developed

6.8.3.2 Pore water pressures around the pile shaft

The excess pore water pressures developed around the pile shaft are shown in Figure 6.118 It can be seen that development of the excess pore water pressures around the pile shaft during a rapid load pile test is complicated as negative pore pressures were developed at some locations whereas at other locations positive excess pore pressures were developed